J911 Studies Peer Reviewers

[Newtons Bit]

I've never really been generally impressed by the folks at the "Journal" of 911 Studies, but my latest encounter with one of their peer reviewers, Tony Szamboti, on this forum leads me to believe that they're not only suffering from group think, but also egregiously incompetent on issues they proclaim to be experts on. Mr. Szamboti posited this (http://forums.randi.org/showpost.php?p=3297371&postcount=167) question to a group of non-engineers:

What does the slenderness ratio of a structural steel column need to be to be in the inelastic buckling range?

What were the slenderness ratios of the central core columns at the collapse initiation sites of the 98th floor in the North Tower and 82nd floor in the South Tower?

I saw this question and jumped into the discussion, posting (http://forums.randi.org/showpost.php?p=3297455&postcount=168):
Inelastic buckling occurs for all slenderness ratios under the Euler limit. That's 4.71 * SQRT(E/Fy). Of course extremely stout members won't buckle inelastically, however none of the columns in the upper floors of the WTC were that stout.

Do you have any clue as to what you're talking about? I recommend picking up an AISC Manual of Steel Construction and see exactly how steel is designed these days. We're not in the 1940's, we know how steel fails now. Maybe you should update you knowledge to modern information.

Mr. Szamboti's replied:

How did I know you would come on.

You used an effective length factor of 1.0 in your letter to Gordon Ross, which is for a pinned connection, when you should have used .5 to .65 for fixed both ends connections for the tower columns. The 1.0 gave you larger slenderness ratios and they still weren't greater than 40. Now you are going to say the tower columns weren't in the short column category and would have been subject to inelastic buckling. The AISC equations you show here and which you used in your paper are conservative for design.

You want to say the tower columns would fail due to buckling. Well how about a test case were an I beam with a slenderness ratio of 20 or lower failed due to inelastic buckling. Do you have any test cases? I have an AISC manual right here. I am familiar with the equations and monograph. You want to go around asking others if they have a clue and you seem to be the one who should be asked that question Mr. Smarty pants.

This is where Mr. Szamboti shows his lack of knowledge as regards to structural engineering. The slenderness ratio that we are talking about, and what Mr. Szamboti struggles to understand, here is defined as K*L/r.

Where:
k = effective length factor
L = length of the column (in)
r = radius of gyratio of the column (in) - [This is a function of the geometric properties of a column]

For the columns that we are talking about, the variables L and r are very well defined and not argued. The effective length factor 'k' is where he slips up. In the commentary of the AISC Manual of Steel Construction (arguably the Bible of how to design steel in structures) this factor 'k' is defined. The first place is in table C2.2 (shown below).

http://forums.randi.org/imagehosting/thum_16329477c5ad0df89f.jpg (http://forums.randi.org/vbimghost.php?do=displayimg&imgid=10058)
Table C-C2.2

Under column (a) it defines the theoretical k value of 0.5 and a recommended design value of 0.65. It's fairly easy to see that this is where Mr. Szamboti thinks the factor k is defined, as the tower exterior columns were moment frames, which means that the top and bottom portions of the columns were fixed. Column (a) shows a column element fixed at the top and bottom, so he used it. And he's very wrong. The commentary clearly explains how this table is to be used on the page before the table, "These range from simple idealizations of single columns such as shown in Table C-C2.2 to complex buckling solutions for specific frames and loading conditions". In his rush to prove me wrong, I can only surmise that he went through the commentary to find an answer to his question, and stumbling upon the first table that seemed to show an answer that confirmed his bias, he lept to a conclusion. An incorrect one not supported by the document he was referencing.

The correct way to calculate this effective length factor is with the nomograph chart, shown below. The nomograph table is for frames which can translate horizontally, this contributes significantly to the stability of the frame.

http://forums.randi.org/imagehosting/thum_16329477c57124fc1f.jpg (http://forums.randi.org/vbimghost.php?do=displayimg&imgid=10057)
Figure C-C2.4

This table looks nonsensical, but it's fairly simple to use. First Ga and Gb need to be defined, which are simply a comparison of the stiffnesses of the columns to the girders. Ga is the comparative stiffness of the top point of the column and Gb is of the bottom. Then, to get the effective length factor, one merely needs to draw a straight line between these two points (see the figure below with Ga = 1.0 and Gb slightly stiffer).

http://forums.randi.org/imagehosting/thum_16329477fcfe5ad3f4.jpg (http://forums.randi.org/vbimghost.php?do=displayimg&imgid=10108)
Example

In this example, the k factor of a frame that has a column stiffness that is roughly equal to the girder stiffness is about 1.4.

It is very easy to see with this table that the lowest factor k that a column in a moment frame can have is 1.0, rendering Mr. Szamboti's statement that it should be 0.5 or 0.65 completely without merit. The purpose of this isn't to belittle or attack Mr. Szamboti as being an incompetent engineer. I'm sure he is an excellent mechanical engineer, however he is not an expert in structural engineering, far from it. He is most definitely unqualified to "peer-review" papers of a structural engineering focus for anyone.

[jhunter1163]

Great post, NB. Even I, a non-engineer, could understand exactly why Tony the Twoofer is wrong.

I've also got one beverage of your choice that says he doesn't answer this criticism.

[uk_dave]

something technical with equations and stuff

And where on youtube can I view this?

[A W Smith]

Dont you mean Monograph (http://forums.randi.org/showpost.php?p=3297646&postcount=170)?


:dl:

[DGM]

It seems lately that they are just trying to sound smart. They don't seem willing to defend the articles just get them out. This has all the signs of a "recruit the uninformed" drive.
It doesn't matter to them if the information is correct as long as they have followers willing to repeat it.

[Newtons Bit]

Great post, NB. Even I, a non-engineer, could understand exactly why Tony the Twoofer is wrong.

I've also got one beverage of your choice that says he doesn't answer this criticism.
He's just quite simply wrong, and it's easy to see how. This is an example where the peer reviewers at J911 can be challenged on their expert status fairly easily. I just hope people like Lisabob are paying attention.

[Coffee]

He's just quite simply wrong, and it's easy to see how. This is an example where the peer reviewers at J911 can be challenged on their expert status fairly easily. I just hope people like Lisabob are paying attention.

Lisabob2 will ignore this. Lisabob2 routinely ignores posts that show the peer review at J911 to be a joke just as s/he ignores anything that go against the words of Jones and Griscom.

[Tony Szamboti]

I've never really been generally impressed by the folks at the "Journal" of 911 Studies, but my latest encounter with one of their peer reviewers, Tony Szamboti, on this forum leads me to believe that they're not only suffering from group think, but also egregiously incompetent on issues they proclaim to be experts on. Mr. Szamboti posited this (http://forums.randi.org/showpost.php?p=3297371&postcount=167) question to a group of non-engineers:



I saw this question and jumped into the discussion, posting (http://forums.randi.org/showpost.php?p=3297455&postcount=168):


Mr. Szamboti's replied:



This is where Mr. Szamboti shows his lack of knowledge as regards to structural engineering. The slenderness ratio that we are talking about, and what Mr. Szamboti struggles to understand, here is defined as K*L/r.

Where:
k = effective length factor
L = length of the column (in)
r = radius of gyratio of the column (in) - [This is a function of the geometric properties of a column]

For the columns that we are talking about, the variables L and r are very well defined and not argued. The effective length factor 'k' is where he slips up. In the commentary of the AISC Manual of Steel Construction (arguably the Bible of how to design steel in structures) this factor 'k' is defined. The first place is in table C2.2 (shown below).

http://forums.randi.org/imagehosting/thum_16329477c5ad0df89f.jpg (http://forums.randi.org/vbimghost.php?do=displayimg&imgid=10058)
Table C-C2.2

Under column (a) it defines the theoretical k value of 0.5 and a recommended design value of 0.65. It's fairly easy to see that this is where Mr. Szamboti thinks the factor k is defined, as the tower exterior columns were moment frames, which means that the top and bottom portions of the columns were fixed. Column (a) shows a column element fixed at the top and bottom, so he used it. And he's very wrong. The commentary clearly explains how this table is to be used on the page before the table, "These range from simple idealizations of single columns such as shown in Table C-C2.2 to complex buckling solutions for specific frames and loading conditions". In his rush to prove me wrong, I can only surmise that he went through the commentary to find an answer to his question, and stumbling upon the first table that seemed to show an answer that confirmed his bias, he lept to a conclusion. An incorrect one not supported by the document he was referencing.

The correct way to calculate this effective length factor is with the nomograph chart, shown below. The nomograph table is for frames which can translate horizontally, this contributes significantly to the stability of the frame.

http://forums.randi.org/imagehosting/thum_16329477c57124fc1f.jpg (http://forums.randi.org/vbimghost.php?do=displayimg&imgid=10057)
Figure C-C2.4

This table looks nonsensical, but it's fairly simple to use. First Ga and Gb need to be defined, which are simply a comparison of the stiffnesses of the columns to the girders. Ga is the comparative stiffness of the top point of the column and Gb is of the bottom. Then, to get the effective length factor, one merely needs to draw a straight line between these two points (see the figure below with Ga = 1.0 and Gb slightly stiffer).

http://forums.randi.org/imagehosting/thum_16329477fcfe5ad3f4.jpg (http://forums.randi.org/vbimghost.php?do=displayimg&imgid=10108)
Example

In this example, the k factor of a frame that has a column stiffness that is roughly equal to the girder stiffness is about 1.4.

It is very easy to see with this table that the lowest factor k that a column in a moment frame can have is 1.0, rendering Mr. Szamboti's statement that it should be 0.5 or 0.65 completely without merit. The purpose of this isn't to belittle or attack Mr. Szamboti as being an incompetent engineer. I'm sure he is an excellent mechanical engineer, however he is not an expert in structural engineering, far from it. He is most definitely unqualified to "peer-review" papers of a structural engineering focus for anyone.


First, you are using the sideways uninhibited nomograph when you should be using the sideways inhibited nomograph. Paragraph 1.8.2 of the Commentary of the AISC manual says connections to floor slabs constitute a braced frame for horizontal stability. This would be considered sideways inhibited and there weren't any columns in the twin towers that did not have floor slabs at their beam connections. Here is a little tutorial on determining K factors for your prematurely cheering fans. Notice how those that are sideways inhibited are less than 1.0.

http://cnx.org/content/m10746/latest/

It sounds like you use the sideways uninhibited nomograph for conservativism.

Oh, and please spare us your canard about mechanical engineers not being structural engineers. What a load of baloney that is and you know it. The structural related courses for mechanical and civil engineers are the same in most schools. Where they diverge is that civils then do more with geotechnical (soils) and transportation (highway building) related courses and mechanicals do more with heat transfer and thermodynamic related courses in the other parts of their curriculums. Buckling can occur in machine elements just as it can with building frames. Mechanical engineers also design the structures for aircraft, automobiles, trains, etc. and buckling is certainly a potential failure mechanism there. There are also dynamic loads involved in these situations which is more complicated.

[Gravy]

Dude doesn't even know the difference between sideways and sidesway.

Which of the columns in your link http://cnx.org/content/m10746/latest/ best represents the exterior columns of the towers, Tony?

Mechanical engineers also design the structures for aircraft, automobiles, trains, etc. and buckling is certainly a potential failure mechanism there. There are also dynamic loads involved in these situations which is more complicated.
You work on antenna design, right?

[Tony Szamboti]

Dude doesn't even know the difference between sideways and sidesway.

Which of the columns in your link http://cnx.org/content/m10746/latest/ best represents the exterior columns of the towers, Tony?


You work on antenna design, right?

Mark, it sounds like the dumbed down version wasn't dumb enough for you.

How about column AB representing the perimeter columns. Do you get it? In fact since they would have been considered fixed on both ends the K factor would have been even lower than the pinned connection shown which has a K factor of .77.

Oh, and you caught a little spelling error. Congratulations. Are you going to try and declare victory with that? You can make a video of how many times you caught spelling errors of people you disagree with.

How did the core columns fail in the towers and what was their K factor? It wasn't Newtons Bit's 1.0 as even the previously blind cheering fans here at JREF must now admit.

[T.A.M.]

I'll soon be expecting someone to stick out their tongue and shout "Nah Nah Nah Nah Nah Nah!"

TAM:)

[Myriad]

First, you are using the sideways uninhibited nomograph when you should be using the sideways inhibited nomograph. Paragraph 1.8.2 of the Commentary of the AISC manual says connections to floor slabs constitute a braced frame for horizontal stability. This would be considered sideways inhibited and there weren't any columns in the twin towers that did not have floor slabs at their beam connections. Here is a little tutorial on determining K factors for your prematurely cheering fans. Notice how those that are sideways inhibited are less than 1.0.

http://cnx.org/content/m10746/latest/

It sounds like you use the sideways uninhibited nomograph for conservativism.


The exercise you reference specifically states that the sidesway inhibited condition of AB, CD, and GF is due to the support at point J. The little triangular thingie in the schematic at point J is a symbol representing a strong immovable anchor point, such as a concrete foundation pier, correct?

What would be the equivalent of point J on the above-ground floors of the World Trade Center?

Seems to me (with all due respect for your expertise) that a WTC tower floor would be more like EH, and the columns more like DE and GH, which are described as sidesway uninhibited in the exercise you referenced. That would agreee with Newton's assessment.

Respectfully,
Myriad

[Tony Szamboti]

The exercise you reference specifically states that the sidesway inhibited condition of AB, CD, and GF is due to the support at point J. The little triangular thingie in the schematic at point J is a symbol representing a strong immovable anchor point, such as a concrete foundation pier, correct?

What would be the equivalent of point J on the above-ground floors of the World Trade Center?

Seems to me (with all due respect for your expertise) that a WTC tower floor would be more like EH, and the columns more like DE and GH, which are described as sidesway uninhibited in the exercise you referenced. That would agreee with Newton's assessment.

Respectfully,
Myriad


You can argue with the AISC manual, which says floor slab connections constitute a braced frame. The reason is obvious and it is the stiffness and great inertia which the floor slabs impart to the connection. Newtons Bit's use of a K factor of 1.0 is conservative for design, as I pointed out earlier, and does not belong in determining what the real K factor was in the failure analysis.

[peteweaver]

Just because the structure below the impact areas of the WTC's were untouched by fire, and undamaged before the collapse, it doesn't mean they were indestructible.
They were capable of supporting the mass of the structure above them, yes, but not, when that mass became dynamic, and fell.

Velocity changes things somewhat.

[Tony Szamboti]

Just because the structure below the impact areas of the WTC's were untouched by fire, and undamaged before the collapse, it doesn't mean they were indestructible.
They were capable of supporting the mass of the structure above them, yes, but not, when that mass became dynamic, and fell.

Velocity changes things somewhat.

You need to show how you get to the dynamic load. I am very skeptical that all of the columns on the initiation floors could have buckled simultaneously, due to fire, and allowed a freefall to create any dynamic load. This is why we are discussing the possibility of buckling and the actual slenderness ratio of the columns plays a big part in that possibility or lack of.

With Gregory Urich's substantiated mass analysis it turns out that the factor of safety of the central core columns was approximately 3.00 to 1 and the perimeter columns were a minimum of 5.00 to 1, when considering gravity loads only. This would need to be overcome to allow a collapse.

[Tony Szamboti]

I'll soon be expecting someone to stick out their tongue and shout "Nah Nah Nah Nah Nah Nah!"

TAM:)

No, that won't happen. However, I would appreciate an apology for being called a moron by a prominent poster here.

[Myriad]

You can argue with the AISC manual, which says floor slab connections constitute a braced frame. The reason is obvious and it is the stiffness and great inertia which the floor slabs impart to the connection. Newtons Bit's use of a K factor of 1.0 is conservative for design, as I pointed out earlier, and does not belong in determining what the real K factor was in the failure analysis.


My copy of the AISC manual (Specification for Structural Steel Buildings, March 9, 2005) does not contain any paragraph 1.8.2, nor does it use the phrase "floor slab" except in only one place, in a section concerned with fire resistance. Thus, I cannot verify your reference to what the AISC manual says about floor slab connections. Please be more specific in the reference (chapter and section name) or quote a sentence or two directly so that I can find the equivalent passage the current manual.

Without a specific valid reference, I can't tell whether the statement you are referencing is correctly interpreted as applying to the types of floors in the WTC.

It seems unlikely that the connections between the columns and floors in the WTC fit the definition of sidesway inhibited, because this would require them (according to table C-C2.2) to be "rotation fixed" when clearly in structural terms they are hinged connections. The floor "slabs" in the WTC towers are not thick enough to provide significant resistance to the rotation of a column connection, even if the columns had been embedded in the concrete rather than, as they actually were, bolted to the ends of the floor trusses.

Furthermore, the towers as a whole seem to far better fit AISC's definition of a moment frame than a braced frame. They did not resemble in any way a "vertical truss system" due to the almost complete lack of diagonal bracing members.

So again, with the exception of the one reference to Commentary paragraph "1.8.2" discussing "floor slabs" that you've offered, neither of which are found in a search of the Commentary, all of the AISC documentation seems to support Newton's view.

Respectfully,
Myriad

[Tony Szamboti]

My copy of the AISC manual (Specification for Structural Steel Buildings, March 9, 2005) does not contain any paragraph 1.8.2, nor does it use the phrase "floor slab" except in only one place, in a section concerned with fire resistance. Thus, I cannot verify your reference to what the AISC manual says about floor slab connections. Please be more specific in the reference (chapter and section name) or quote a sentence or two directly so that I can find the equivalent passage the current manual.

Without a specific valid reference, I can't tell whether the statement you are referencing is correctly interpreted as applying to the types of floors in the WTC.

It seems unlikely that the connections between the columns and floors in the WTC fit the definition of sidesway inhibited, because this would require them (according to table C-C2.2) to be "rotation fixed" when clearly in structural terms they are hinged connections. The floor "slabs" in the WTC towers are not thick enough to provide significant resistance to the rotation of a column connection, even if the columns had been embedded in the concrete rather than, as they actually were, bolted to the ends of the floor trusses.

Furthermore, the towers as a whole seem to far better fit AISC's definition of a moment frame than a braced frame. They did not resemble in any way a "vertical truss system" due to the almost complete lack of diagonal bracing members.

So again, with the exception of the one reference to Commentary paragraph "1.8.2" discussing "floor slabs" that you've offered, neither of which are found in a search of the Commentary, all of the AISC documentation seems to support Newton's view.

Respectfully,
Myriad

I have the Eighth edition of the AISC manual and the Chapter 5 Commentary was effective 11/1/1978. Maybe they changed the paragraph number in later editions. Chapter 5 Section 1.8 of the Eighth edition is titled "Stability and Slenderness Ratios". You should look for a paragraph with that title.

I think you need to do an analysis to show that any of the columns in the towers would have performed as pinned connections rather than fixed on both ends. Diagonal bracing is not the only thing that causes a frame to act as braced.

[rwguinn]

I have the Eighth edition of the AISC manual and the Chapter 5 Commentary was effective 11/1/1978. Maybe they changed the paragraph number in later editions. Chapter 5 Section 1.8 of the Eighth edition is titled "Stability and Slenderness Ratios". You should look for a paragraph with that title.

I think you need to do an analysis to show that any of the columns in the towers would have performed as pinned connections rather than fixed on both ends. Diagonal bracing is not the only thing that causes a frame to act as braced.

I do not have access right now to the manual, and seldom if ever use it in my job.
You are right that diagonal bracing is not the only thing that will cause it to act as braced--and the intact structure would act that way--outside panels acting as shear panels between verticals.
But the tower was not intact. The bracing (shear panels) had been compromised--breached, even.
The floors do not restrain the verticals from rotation, however, and the top of a set of columns is assuredly not fixed. The whole floor is able to translate WRT the floor below, especially when the shear panels go away.

Also you are using "conservative" in a way I am unfamiliar with.
A conservative design, to me, means designed to the worst-possible scenario--and a conservative analysis means the worst possible load case with the worst possible boundary conditions--or the most likely to cause failure. You seem to be implying it is something opposite that...

[Newtons Bit]

First, you are using the sideways uninhibited nomograph when you should be using the sideways inhibited nomograph. Paragraph 1.8.2 of the Commentary of the AISC manual says connections to floor slabs constitute a braced frame for horizontal stability. This would be considered sideways inhibited and there weren't any columns in the twin towers that did not have floor slabs at their beam connections. Here is a little tutorial on determining K factors for your prematurely cheering fans. Notice how those that are sideways inhibited are less than 1.0.

http://cnx.org/content/m10746/latest/

It sounds like you use the sideways uninhibited nomograph for conservativism.

Oh, and please spare us your canard about mechanical engineers not being structural engineers. What a load of baloney that is and you know it. The structural related courses for mechanical and civil engineers are the same in most schools. Where they diverge is that civils then do more with geotechnical (soils) and transportation (highway building) related courses and mechanicals do more with heat transfer and thermodynamic related courses in the other parts of their curriculums. Buckling can occur in machine elements just as it can with building frames. Mechanical engineers also design the structures for aircraft, automobiles, trains, etc. and buckling is certainly a potential failure mechanism there. There are also dynamic loads involved in these situations which is more complicated.

First, download and read the commentary for AISC-360-05 (http://www.aisc.org/Content/ContentGroups/Documents/2005_Specification/Commentary_2005_Specification_FINAL_revision.pdf). At least this way we can be referencing the same document. Then, admit you've made a mistake because you've yet again failed completely to understand what you're talking about. It's a simple thing to do.

Let's look at the example you've provided. The author provides a diagram representing a building section:

http://forums.randi.org/imagehosting/thum_163294781014125ed2.jpg (http://forums.randi.org/vbimghost.php?do=displayimg&imgid=10135)

The important thing to note is that point J is a pinned connection. The author is correct in saying that sidesway is inhibited because point J, and thus the beams connected to it, cannot translate. Point J represents a stiff lateral element, a braced frame or a shear wall perhaps. It does not represent a moment frame.

The author then shows how to calculate k for the top portion of the building, columns DE and GH. These he says are, "there is no sideways bracing for the top portion of the frame". That's because they are not connected to a braced frame or a shear wall but rather a moment frame (or a cantilevered column, it depends on what the connections look like).

[Tony Szamboti]

Great post, NB. Even I, a non-engineer, could understand exactly why Tony the Twoofer is wrong.

I've also got one beverage of your choice that says he doesn't answer this criticism.

Your use of childish ad hominem along with unwarranted confidence did not add anything to this discussion. It turns out you owe someone a beverage.

[Newtons Bit]

You can argue with the AISC manual, which says floor slab connections constitute a braced frame. The reason is obvious and it is the stiffness and great inertia which the floor slabs impart to the connection. Newtons Bit's use of a K factor of 1.0 is conservative for design, as I pointed out earlier, and does not belong in determining what the real K factor was in the failure analysis.

No it doesn't. Floor slabs brace beams against buckling (both lateral torsional and compressive), a braced frame a slab does not make.

Stop making things up.

[Tony Szamboti]

First, download and read the commentary for AISC-360-05 (http://www.aisc.org/Content/ContentGroups/Documents/2005_Specification/Commentary_2005_Specification_FINAL_revision.pdf). At least this way we can be referencing the same document. Then, admit you've made a mistake because you've yet again failed completely to understand what you're talking about. It's a simple thing to do.

Let's look at the example you've provided. The author provides a diagram representing a building section:

http://forums.randi.org/imagehosting/thum_163294781014125ed2.jpg (http://forums.randi.org/vbimghost.php?do=displayimg&imgid=10135)

The important thing to note is that point J is a pinned connection. The author is correct in saying that sidesway is inhibited because point J, and thus the beams connected to it, cannot translate. Point J represents a stiff lateral element, a braced frame or a shear wall perhaps. It does not represent a moment frame.

The author then shows how to calculate k for the top portion of the building, columns DE and GH. These he says are, "there is no sideways bracing for the top portion of the frame". That's because they are not connected to a braced frame or a shear wall but rather a moment frame (or a cantilevered column, it depends on what the connections look like).


You can try to twist it all you want. The reality is that the columns in the Twin Towers would not have behaved as pinned connections, as you try to claim. They would have acted as being fixed at both ends, and sidesway inhibited, due to the stiffness and inertia provided by the composite action of the floor slabs fastened to beams, in both the core, and the perimeter. I intend to show this and that you are wrong. Your paper will get a real peer review and it will be shown that you are using conservative design values for the slenderness ratios, instead of actuals, to support your contention that the columns could have easily buckled.

This post will be my last here on the subject.

[CHF]

This post will be my last here on the subject.

I can see why, Tony.

How about doing your cause a favour by submitting your "research" for peer-review in a REAL scientific or engineering publication.

[twinstead]

Your paper will get a real peer review and it will be shown that you are using conservative design values for the slenderness ratios, instead of actuals, to support your contention that the columns could have easily buckled.


I shudder to think what would happen to your work if it ever gets a 'real' peer review.

[Newtons Bit]

You can try to twist it all you want. The reality is that the columns in the Twin Towers would not have behaved as pinned connections, as you try to claim. They would have acted as being fixed at both ends, and sidesway inhibited, due to the stiffness and inertia provided by the composite action of the floor slabs fastened to beams, in both the core, and the perimeter. I intend to show this and that you are wrong. Your paper will get a real peer review and it will be shown that you are using conservative design values for the slenderness ratios, instead of actuals, to support your contention that the columns could have easily buckled.

This post will be my last here on the subject.

I never said they were pinned, I said they were MOMENT frames. That's sort of the opposite of pinned in case you didn't know. Don't put your failure to understand simple engineering on me. Even the example you linked to showed you're wrong and you twist and twist and twist to try to get it to say you're right. And when confronted about it, you run away.

And how will my "paper" get a real peer-review? Are you going to submit it under false pretenses? That's not very nice, Mr. Szamboti.

[Myriad]

I have the Eighth edition of the AISC manual and the Chapter 5 Commentary was effective 11/1/1978. Maybe they changed the paragraph number in later editions. Chapter 5 Section 1.8 of the Eighth edition is titled "Stability and Slenderness Ratios". You should look for a paragraph with that title.


Thank you for the additional information. It's clear that there is no single corresponding paragraph in the 2005 edition. The relevant Commentary chapter is now chapter C (chapters are lettered and appendices are numbered), "Stability Analysis and Design," and section C2 within that chapter discusses methods of caclulating required strengths, including "traditional" methods that use the Effective Length Factor K.

I see nothing in that chapter or elsewhere in the current manual that suggests that the WTC towers are any exception of the general rule that K values in braced frames as well as moment frames are normally greater than or equal to 1.0. (see paragraphs C1.3b, C1.3c). There is mention (in the text accompanying the C-C2.3 nomograph) that column ends "rigidly attached to a properly designed footing" (emphasis added) can be taken to have relatively low values of G, which would lead to relatively lower values of K. However, the above-ground floors in the towers, despite being partially constructed of concrete, were in no way "footings" and in any case the column ends were not (very) rigidly attached to them.

More generally, there is nothing that suggests that the general nature of the horizontal connections between column ends, such as their total mass, their length, the materials they're made of, or whether or not they happen to be used as floors, has any bearing (no pun intended) on the effective value of K for the columns. (Naturally, the structural properties of the horizontal members would affect other aspects of any stability analysis, but they don't figure into the K of the columns.)

I think you need to do an analysis to show that any of the columns in the towers would have performed as pinned connections rather than fixed on both ends. Diagonal bracing is not the only thing that causes a frame to act as braced.


"Fixed at both ends" doesn't imply K < 1.0. Actually, according to AISC, braced-frame systems are analyzed and designed as pin-connected systems, with a K of 1.0. Moment-frame systems, which better describes the WTC towers, generally use K values greater than 1.0:

Moment-frame systems rely primarily upon the flexural stiffness of the connected beams and columns, although the reduction in the stiffness due to shear deformations can be important and should be considered where column bays are short and/or members are deep. Except as noted in Section C2.2a(4), Section C2.2b and Appendix 7, the design of all columns and beam-columns must be based on an effective length, KL, greater than the actual length determined as specified in Section C2.(emphasis added)

Lest you get the impression that Sections C2.2a(4), C2.2b, and Appendix 7 contain a statement that these rules don't apply if the columns in question happen to be connected to floors (aren't most columns in most buildings connected to floors?), the quote continues:

The Direct Analysis Method in Appendix 7. as well as the provision of Sections C2.2a(4) and C2.2b, provide the means for proportioning columns with K = 1.0.


In other words, those sections cover alternate "direct" analysis methods in which K is not used as a variable; instead the various influences of the physical parameters that figure into K are accounted for separately (generally, in a computer model).

So far, all the information from the AISC manual appears to support Newton's assessment.

Respectfully,
Myriad

ETA: Cross-posted with the last several posts by Newton, Tony, and others. It seems that learning the principles of structural engineering from reference sources can take some time. Who'd'a thought?

[Jonnyclueless]

Your paper will get a real peer review and it will be shown that you are using conservative design values for the slenderness ratios, instead of actuals, to support your contention that the columns could have easily buckled.

This post will be my last here on the subject.

Uh oh, does this mean Charlie Sheen will be reviewing it? Maybe Webster?

[jhunter1163]

Well, it seems I was wrong about Tony's response (or lack thereof). Since I'm nothing if not a gentleman of my word, I'll buy Tony a beverage of his choice. And NB too, for that matter.

However, Tony is still wrong.

[Newtons Bit]

ETA: Cross-posted with the last several posts by Newton, Tony, and others. It seems that learning the principles of structural engineering from reference sources can take some time. Who'd'a thought?

You seemd to have it figured out with no problems. It's pretty straight forward.

[Gravy]

Aww, I was looking forward to toying with the twoofer some more, but I had to go to bed. It's bad when a tour guide can immediately see why Szamboti, who claims to be a mechanical engineer, doesn't know what he's talking about.

Suppose I just made everything up on my tours. Wouldn't that be wrong of me?

[R.Mackey]

And how will my "paper" get a real peer-review? Are you going to submit it under false pretenses? That's not very nice, Mr. Szamboti.

We should be on the lookout for this... it wouldn't be the first time, as I recall. Truly those at the "Journal" for 9/11 Studies have a demented concept of peer-review!

[Newtons Bit]

I think we can officially call this closed. I'm thinking it might be useful to compile a list of errors that the peer-reviewers of the "journal" make so as to challenge their credibility.

I'm tired of truthers calling them experts on buildings. We need a big long list of links that show they're not. And while just linking to the papers on the "journal" would suffice for most rational folks to see they're B.S. I'd like to see something more substantial where it's easy to focus on just one point and show how they're completely wrong. Especially points where the truther himself won't admit he's wrong when even a tour guide can understand the math involved to see he is.

[JamesB]

We should be on the lookout for this... it wouldn't be the first time, as I recall. Truly those at the "Journal" for 9/11 Studies have a demented concept of peer-review!

No kidding, I write an e-mail to a psychology PhD candidate, and the next thing I know it was a "peer reviewed" paper.

[AZCat]

I think we can officially call this closed. I'm thinking it might be useful to compile a list of errors that the peer-reviewers of the "journal" make so as to challenge their credibility.

I'm tired of truthers calling them experts on buildings. We need a big long list of links that show they're not. And while just linking to the papers on the "journal" would suffice for most rational folks to see they're B.S. I'd like to see something more substantial where it's easy to focus on just one point and show how they're completely wrong. Especially points where the truther himself won't admit he's wrong when even a tour guide can understand the math involved to see he is.

Thanks to you (and the others) for taking the time to explain in detail the calculations for determining slenderness ratios for columns. I had this back in school but not as in-depth (went for the thermo track).

[Newtons Bit]

Word of advice to Mr. Szamboti: when inventing things on the internet, make sure that the people reading it don't have access to what you are referencing.

Mr. Szamboti stated that:
Paragraph 1.8.2 of the Commentary of the AISC manual says connections to floor slabs constitute a braced frame for horizontal stability.

That's pretty specific. Unfortunately for him, I have access to an AISC 8th edition that has the spec he's quoting. And I've attached the page from the spec below. It doesn't say what he says it does. Specifically it states:

In trusses and in those frames where lateral stability is provide by adequate attachment to diagonal bracing, to shear walls, to an adjacent structure having adequate lateral stability, or to floor slabs or roof decks secured horizontally by walls or bracing systems parallel to the plane of the frame, the effective length factor, K, for the compression members shall be taken as unity, unless analysis shows that a smaller value may be used.

Bolding mine. What that is saying is that floor slabs or roof decks that connect to a gravity column AND a brace constitute a braced frame. This is pretty much common sense. The braced frame provides the lateral system and the gravity column just goes along for the ride.

The very next section in the book that Mr. Szamboti was referencing defines unbraced frames. It is very clear that the WTC exterior columns are in this category AND that K shall not be less than 1.

In frames where lateral stability is dependent upon the bending stiffness of rigidly connect beams and columns, the effective length KL of compression members shall be determined by a rational method and shall not be less than the actual unbraced length.

I think we have ourselves a charlatan.

[ref]

Thanks NB :) good work.

[beachnut]

very nice work -

[NicoS]

Hello everyone.
(my first post in this forum...)

Didn't NIST compute a model of the towers and run a simulation to describe the very first instants of the collapse initiation? (I think it was WTC2, as far as I remember) They shall have choosen a K factor, or they shall have choosen a model to represent the structure (including the columns to beams connections) in whitch K can be calculated. What about this K factor in those simulations?
And more generally, isn't it obvious in those simulations that buckling occured?(it is an opened question, I'm not familiar enough with these NIST simulation to make it afirmative).

Thanks for your attention.
And sorry for my english speaking, I'm french (nobody's perfect)